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Page 1: 8.6.3 Buoyancy and weight forces

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FL = CL An γw (u2/2g) (8.1)

Where

CL = empirical lift coefficient

An = projected area of solid body normal to flow

γw = specific weight of water

g = gravitational acceleration

u = magnitude of flow velocity

Based on the Reynolds number of the flow regime and information from the literature,

the lift coefficient for the present situation would likely be in the range of 0.3 to 1.

Substituting in to the equation above (An = 1.2 m2, u = 1 ms-1, ρw = 1024 kg/m3) and

using a range of CL values (from 0.3 to 1.0) yields a lift force of between 185 and 614

N/m of pipe length.

8.6.3 Buoyancy and weight forces

The buoyancy of the pipeline per metre of pipe length should also be considered. In

general the buoyancy force is equal to the weight of the fluid displaced. This is

expressed as:

FB = γw V (8.2)

γw is the specific weight of water and V is the volume of water displaced.

For an empty pipe with an outside diameter of 1.2 m, the buoyant force is equal to 45.2

kN per meter of pipe length. This will be reduced by the weight of the pipe but the wall

thickness is not known at this stage. The buoyant force is in opposition to the weight of

the fluid being carried inside the pipe (in this case wastewater) and the weight of the

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pipe which both act vertically downward. Since the weight of the wastewater will be a

function of the interior diameter (ID) of the pipe and because wastewater should be

slightly less dense than sea water, the net force (excluding the weight of the pipe) will

be upwards, without the use of a concrete weight coat on the pipe. The final

magnitude of the buoyant or gravitational force will then be dependent on the pipe

material, its weight and wall thickness, but the excess buoyancy will need to be offset

by a concrete weight coat.

8.6.4 Scour

Generally some amount of scour is a desirable end result for a pipeline laid on the sea

bed. However, if the pipeline is laid across bed material of differing erodibility, the result

may be a free span of pipe that is not in contact with the seabed. This is a potentially

dangerous situation that could lead to pipeline failure. In general, the Coastal

Construction Manual recommends a simple relationship between maximum scour

depth and pipeline diameter for situations where the current velocity exceeds the

critical velocity for bed erosion. This relationship is given as:

Sm = 0.7 D (8.3)

where Sm is the maximum scour depth in metre and D is the pipeline diameter in

metres.

8.6.5 Wave impact

The most likely locations where wave forces or scour may be come problematic are at

the locations where the pipeline emerges from the estuary onto dry land or in the

intertidal zone. In these locations and under certain circumstances, the wave heights

could become a significant fraction of the water depth and either break or generate

significant impact forces on the pipeline. For the case of wave impact, the force

exerted on the structure can be expressed as:

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FU = CU Az γw (w2/2g) (8.4)

Where

CU = laboratory derived slamming coefficient

Az = projected area of solid body in the horizontal plane

w = vertical component of flow velocity at the level of the object

γw = specific weight of water

g = gravitational acceleration

The Coastal Construction Manual cites values of CU acting on horizontally mounted

cylinders (i.e. a pipeline) to be in the range of 4.1 to 6.4. Under the range of conditions

expected at the site (wave height = 0.2 m, period = 2 sec) vertical velocities could

range up to 0.45 m/s. Assuming a CU of 5 and the quantities listed above, the force

acting on the pipeline would be approximately 622 N per meter of pipe length.

Additional scour protection and reinforcement might be considered for the section of

the pipe that might fall in these conditions.

8.6.6 Free spans

Due to differential bed erosion or bathymetric highs and lows, the pipeline may end up

with sections that are not directly supported by the seafloor in a free span. Free spans

can be problematic for two reasons. First, the pipeline must be analyzed for the

additional stresses that include the bending moment and tensile forces that will be

generated at the mid-point of the free span. Second, fluid flow due to currents and

waves can lead to oscillatory motion of the pipe. ‘Lift’ forces can be generated in a

downward direction if there is a narrow space under the pipe where water can flow. If

the flow is oscillatory (i.e. from waves) then alternating positive and negative ‘lift’ can

be generated resulting in a vibrational motion that can lead to failure. The frequency of

this flow-induced oscillatory motion should be de-tuned from the frequency of the free

span. This level of analysis requires information on the pipe material to be used and

cannot be done here. Overall, it is preferable for the pipeline to be laid at a depth where

risk of free spanning is minimised.

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8.7 Summary and Recommendations

Two submarine harbour-crossing pipeline route options have been assessed for the

worst case assumption that the pipeline is, or becomes, exposed. Both routes are

predicted to alter the hydrodynamics and sediment transport in the main channel, but

only at a very local scale. However, the northern route near the railway bridge (Option

5B) is found to be superior in relation to flow alterations, sedimentary impacts and

visual appearance. One negative influence of the southern pipe (Option 5C) is the

identified potential for muds to build up in the zone between the pipe and railway

causeway. These muds may degrade the inter-tidal zone, change the biota and

potentially lead to polluted mud contamination in the future. The long crossing over the

inter-tidal sand flat could also “weaken” the natural bonds in the harbour bed

sediments. The southern route is a difficult traverse over multiple channels and sand

banks, and could interfere with boating safety.

The northern route (Option 5B) with the pipeline placed along the existing railway

causeway is recommended for these reasons and because negligible environmental

effects are anticipated along this route, noting that the biological effects of direct burial

of the seabed by the widened causeway is not considered in this report. The beach

crossings at the western and eastern ends of the pipeline are considered in Chapter

10.

8.8 Shoreline traverse of the proposed pipeline

In the next chapter, the effects of a proposed walkway to cover the pipe as it traverses

south at the top of the beach on the western side of the channel (Figure 8.15) is

considered. Likely physical and social impacts of this route include:

• Build up of muds between the pipe and the natural shoreline

• Changed wave and current climate along the shoreline

• Changed natural character of the existing shoreline

• Difficulties associated with water access for the public

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• Potential boating/anchoring hazard at the northern extremity

Figure 8.15. Proposed pipeline routes

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9 Walkway along the western shoreline beach

9.1 Introduction

URS provided the latest design for the proposed western shoreline walkway, which

will serve the dual purpose of a promenade and a structure to cover the pipeline as

it traverses north along the upper section of beach towards the railway bridge

(Figure 9.1). The walkway is to be constructed of rock as an embankment. In

addition, shoreline sand nourishment (some 3000 m3) is proposed for the beaches

along the western side and this was incorporated into the bathymetry grids.

Numerical modelling was undertaken to determine the secondary effects of the

proposed walkway on the channel circulation and sand banks. The effects on the

upper beach or the direct effect of burial by the structure are not considered.

Specifically, the questions to be addressed are:

• what effect will the walkway have on the harbour circulation and on existing

sand banks, particularly the sand bank to the north between First Ave and

the railway bridge (Figure 9.2); and

• will there be an impact on the restaurant to the immediate north of the bridge

(Figure 9.3)

• will there be toe scour at the foot of the walkway structure and what are the

implications for structure stability

A fine scale numerical model of the region was developed to predict the changes to

currents and inferences are drawn in relation to the likely impacts on the sediment

transport. Sediment transport modeling to determine the long-term effects or the rate

of scour along the toe of the walkway was not undertaken.

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9.2 Numerical modeling bathymetry

The bathymetry grids, “nourished beaches” and “walkway”, were created by

incorporating new shoreline surveys provided by URS. For completeness, both (1)

nourished beach and (2) nourishment plus the walkway were modeled separately,

but the differences are very small and no substantial difference in the outputs was

expected. The existing model grids were updated with the surveyed cross-shore

transects to create the bathymetry shown in Figures 2.11 and 2.12, noting that the

multi-beam survey is not able to measure in shallow inter-tidal and sub-aerial zones,

but that ASR undertook an inter-tidal survey over this region for the project (Figure

2.7).

As such, the bathymetry in these simulations may be somewhat improved on the

western shoreline, but the results indicate that changes to the currents were

insubstantial. The model simulations use the same boundary conditions, as in

previous model runs because of the small bathymetric changes.

9.3 Modelling results

Figures 9.4 and 9.5 show peak flood and ebb flow respectively. Figure 9.6 shows

the residual currents. As the velocity changes are small, the difference in the

residual velocities between the existing case and with the walkway included is the

most informative output because it averages all currents over the 15-day modeled

period (Figure 9.7). The residual velocity differences (Figure 9.7) show that:

• The residual velocities change by 1-2 cm/s within 50-70 m of the walkway

embankment

• Currents become more flood dominant along the walkway. This change is

small, but sediment modeling would be needed to provide a full assessment

of the impact. From the modeling, the likely effect is minimal, acknowledging

that the new residual current sometimes opposes the movement of sand

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from the inter-tidal flat along the western shoreline which evidently forms the

coastal nearshore sand body along this shoreline (see Figure 2.15 and

further discussion below).

• With the exception of the locations noted above, currents are unchanged

Figure 9.1. Location map of the western shoreline beach

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Figure 9.2. Section of the proposed walkway from First Avenue to The Strand.

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Figure 9.3: Western end of the Railway bridge with the car park shoreline works and the

restaurant in the background.

9.4 Discussion

The walkway is discussed in 4 regions which are distinguished as follows:

Southern end to 4th Avenue: The inter-tidal zone is wide in this region and currents

near the toe of the walkway are slow. ASR did not undertake studies of inter-tidal

beach erosion at the foot of the walkway. However, further offshore, it would be

expected that the sand flat will not be affected by the presence of the walkway.

4th Avenue to 1st Avenue: The current intensities increase in this area near the

shoreline and the inter-tidal zone is narrow. The walkway projects further out into the

channel at First Avenue and the toe of the walkway is sub-tidal at all tides. The

expectation is for some erosion at the toe of the walkway, particularly towards First

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Avenue. It is recommended that sheet piling be adopted along this section, and

monitoring to discern if erosion depths prove to be greater. If necessary, a top up of the

sand or scour protection could be adopted.

1st Avenue to the Railway Bridge: The currents are fast in this area and the walkway

is adjacent to the channel with segments of the toe being sub-tidal at all tides. The

section is in the lee of the artificial “headland” created by the walkway near First

Avenue. Here the walkway covers the narrow beach and extends to the edge of the

Town Reach channel. It would be expected that the beach in front of the walkway will

be lost slowly by erosion. Additional foundations for the walkway embankment may be

needed, such as sheet piling. Erosion at the toe could be exacerbated by the general

retreat of the shoreline along this zone, as shown by the aerial photographs.

The shoreline region to the north of First Avenue appears from the bathymetry to be a

continuation of the broader inter-tidal zone and beach to the south (Figure 9.8). Thus,

burial of this sediment pathway under the walkway embankment would disrupt this

natural sand flow and potentially lead to some erosion along the toe of the walkway

embankment.

It is recommended that sheet piling be adopted along this section, and monitoring to

discern if erosion depths prove to be greater. If necessary, a top up of the sand or

scour protection could be adopted.

North of the railway bridge in the vicinity of the restaurant: No changes to the

currents were identified. Figure 9.9 shows the modelled flow velocities for a peak flood

tide. The model shows low flow velocities of around 0.1 m/s, indicating that the

location is not prone to high sediment loads. Inspection of aerial photographs of the

region at low tide shows a sand bar which appears to be inactive (Figure 9.2) and the

sand grain sizes are also coarser here than in the main channel (Figure 2.18). Thus it

is likely that this region will be relatively stable through time, with a slow adjustment to

the new circulation conditions. In accordance with these conclusions, slow erosion of

the shoreline to the immediate south of the bridge (in front of the car park) has been

observed (Figure 9.3).

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9.5 Anticipated scour depths

From the modelling undertaken and the bathymetry surveys, it would appear that the

beach and sub-tidal sand body around 1st Avenue is a continuation of the large inter-

tidal sand flat to the south (Figure 2.15). The residual currents on the sand flat are

directed to the north (Figure 4.10) and the wave fetches are largest from the south

which may be anticipated to slowly drive sediment north along the beach.

The beach sediments are coarser than those in the channel (of the order of 1.0 mm

versus 0.4 mm, Figure 2.18) while the sediments that have deposited in front of the car

park to the immediate south of the railway bridge are also coarse, around 1.0 mm

(Figure 2.18). Thus, this would imply that the sands have been swept along the beach

to be deposited in the sand bar near the bridge.

By placing the proposed walkway embankment along this section of beach and

shoreline, some of this sediment pathway will be buried, leaving only the deeper sub-

tidal region available for continued sediment transport. Other factors that may influence

the outcomes are the proposed boatramps along this section which will act as small

groynes.

Thus, it may be anticipated that the disruption of the shoreline sediment route could

lead to some scour of the seabed along the toe of the embankment.

Only limited information is available to determine the likely rate and final depth of the

scour, but the historical surveys north of the bridge in the Town Reach are instructive.

In this region, the shoreline has been similarly walled, and the original beach has been

buried. Figure 2.34 along Transect 3 shows a continuous trend for the seabed along

the western shoreline to erode. The erosion appears to be mirrored by a compensating

accretion on the other side of the deeper section of the channel. While the historical

survey comparisons may contain some error and the situation at 1st Avenue is not

exactly the same, the transects nevertheless show the expected trend. The rate of

erosion appears to be in the range of 2-4 m over a period of approximately 100 years.

At 1st Avenue, the sediment supply may remain partially intact with the large inter-tidal

sand flat and the channels to the south unchanged by the works. In addition, the

currents are less at around 0.6-0.7 m.s-1 offshore of 1st Avenue and 0.9 m.s-1 through

the transect location (Figure 9.5). Accordingly, it would be reasonable to estimate the

scour potential around 1st Avenue to be no worse than half that on the transect. Thus, it

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is anticipated that scour at the toe of the walkway may be in the range of 1-2 m over a

period of decades.

Figure 9.4. Peak flood velocities for the nourished beach” and “walkway” bathymetry.

Figure 9.5. Peak ebb velocities for nourished beach” and “walkway” bathymetry.

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Figure 9.6. Residual velocity vector averages over a 15 day spring-neap cycle for the nourished beach” and “walkway” bathymetry with zoomed image over the area of interest.

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Figure 9.7. The difference in residual velocity vector averaged currents. Hydrodynamic variations are limited to the local region were bathymetric changes exist.

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Figure 9.8. Fine grid (2 m) model bathymetry for Tauranga Harbour in the vicinity of the new bridge and the western bank.

Figure 9.9. Modelled flow velocities for a peak flood tide. Note the lower flow velocities near the western wall.

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9.6 Conclusions and Recommendations

Numerical modelling undertaken to determine the effects of the proposed walkway

embankment on the channel circulation and sand banks suggest that the residual

velocities change by 1-2 cm/s within 50-70 m offshore of the walkway embankment

while currents are unchanged elsewhere. From the southern end to 4th Avenue, the

offshore, sand flat is unlikely to be affected by the presence of the walkway. However,

along the sector from 4th Avenue to the railway bridge, the expectation is for some

erosion at the toe of the walkway, particularly around First Avenue. Additional

foundations for the walkway embankment may be needed along this section, such as

sheet piling.

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10 Burial of the pipeline

10.1 Introduction

Here, a brief assessment of the effects during construction of burying the submarine

pipeline is undertaken. The key issues to consider were:

• Depth of burial

• Formation of sediment plumes during construction

• Coffer dam sediments

The following methodology was proposed by URS for installing the submarine pipeline

adjacent to the existing railway bridge:

1. Pipe is laid across the harbour bed surface in one piece.

2. A jetting machine is bolted around the pipe at one side of the harbour, and

travels along the pipe to the other end, jetting out the harbour bed material as it

moves.

3. The depth of jetting may vary from 200 mm to 400 mm per pass.

4. The jetting machine travels back and forth until the 2 pipes at 6 m spacing are

laid under the harbour bed. Cover of 1 m has been allowed for, so the jetted

trench will be about 2.1 m deep in total.

5. Crossing length is approximately 600 m when the sledge movement is typically

1 m/minute.

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6. There will be an area of approximately 30 m width from the centreline of the two

pipes (namely 12 m either side of each pipe) that will be totally disrupted by the

excavation process.

The process of excavation would involve creation of slurry with about 15% solids and

85% fluid. The sediments in the region have up to 10% mud content and so the muds

would likely disperse into the water column to create a visible plume during the

construction operation.

10.2 Depth of burial of the pipe

Bathymetry transects comparisons in the Town Reach show natural fluctuations in the

channel depths of 1-2 m over the last 100 years (Figure 2.34), which is the design life

of the submarine pipeline. Fluctuations appear to be greater than 2 m on the steeper

channel flanks. Thus, if the pipeline is buried 1 m below the current bed level, as

planned, there is some likelihood that sections of the pipeline may become exposed

over its design life. Without repetitive historical bathymetry data, the probability of

exposure cannot be precisely determined. However, if the pipeline is to remain buried

throughout its design life, a burial depth of at least 2 m would be in accordance with the

fluctuations identified by the channel cross-section analysis.

10.3 Sediment plumes during construction

Suspended sediment plume modeling may be required to predict the likely effects of

turbidity arising from the construction operation. Sediment plume modeling is very

similar to modeling undertaken for dredging when sediment is placed into suspension

by the dredging operation. In the model, the bed material being suspended by the

trenching equipment is simulated. The full distribution of suspended grain sizes is

broken into 10-20 fractions. Each fraction then falls to the bottom with a velocity

dependent on the grain size. The turbulence in the water column associated with the

currents helps to oppose the fall velocity and keeps the material in suspension for

longer than in still water. These processes are included in the model which predicts:

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• Concentrations in the water column

• The patterns and locations where sediment drops to the seabed with contours

of grain impact distribution

• Depth changes associated with the deposition and full contouring of the

changes in bed levels

This combination of outputs allows forecasts to be made of biological impacts, the likely

changes to water quality, the spread of the plume and the changes in bed level around

the operation.

Prior to initiating any modeling, a desktop analysis is presented here to determine if the

plume will be visible or high concentration.

We have conservatively estimated the volume of sand+silt that will be put into

suspension by the jetting machine. This is given below (depending on the depth of

each pass).

• For the 200 mm scenario:

40,000 m3 of bed sediment material will be displaced in total assuming two pipes.

11 passes will be needed at 3.3 hours for each pass.

• For the 400 mm scenario:

26,000 m3 of bed sediment material will be displaced in total assuming two pipes.

6 passes will be needed at 3.3 hours for each pass.

Some material may fall back into the hole, or immediately to the side of the hole, and

so some further discussion about the operation of the equipment will be needed. It was

subsequently determined by URS after discussions with contractors that the volume of

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sediment disturbed will be less than 20,000 m3 and so the analysis here is potentially

over-estimating the plume intensity.

If 20,000-40,000 m3 was deposited over a region some 100 m long and across the full

channel (assuming 400 m width), the depth of deposition would exceed 0.5-1.0 m over

the full region.

Consideration of natural background levels in the channel using data from EBOP

indicate that, around the railway bridge, the concentrations of TSS in the water column

are:

Table 10.1: Background concentrations of TSS (Total Suspended Solids) in Tauranga Harbour (information provided by URS).

Concentration measure

Concentration (mg/L)

Minimum 3.3

Maximum 355

Average 25.6

Median 16

Consequently, a plume due to the construction having a concentration above the

average background level of 25.6 mg/l would be clearly visible.

Mud Discharge

Assuming the worst case scenario, 40,000 m3 displaced material over 11 passes, this

equates to 3640 m3/pass. At an assumed mud fraction of 10%, we now have 364 m3 of

plume forming muds entrained per pass. At 3.3 hours per pass and using a density of

1432 kg/m3 for mud, this equates to a discharge of 43.8 kg (4.4 x 107 mg) of muds per

second.

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Volume Flow Rate

During a tidal cycle, the volumetric flow rate for the entire cross section at the bridge

crossing is between 100 and 1000 m3/sec (105 to 106 L/sec) depending on the stage of

the tide.

Concentrations

Using the gross amount of mud discharged divided by the gross flow rate yields overall

average concentrations of between 440 mg/L (low flow) and 44 mg/L (high flow). Given

that the muds will be discharged more as a point source at the location of the

excavator, the actual concentrations around the equipment would be higher than the

concentrations averaged across the full cross section presented here. It can be

concluded therefore that the plume will be visible and could frequently exceed the

background levels.

For a more detailed understanding of the plume and the settlement locations of the

displaced sediments, it would be necessary to undertake a sediment settlement

analysis using the calibrated numerical models.

10.4 Coffer dams

For the submarine pipeline option, the construction method calls for excavation

protected by a coffer dam at the points on both sides of the channel where the pipeline

enters the harbour seabed. URS has expressed concern that the excavated sediment

could wash away, spread across a broader region or cause localized problems with

currents.

To overcome the problems of sediment loss and broader effects across the region, the

sediment taken out for the coffer dams could be pumped into geotextile mega-

containers. These would store the sediments safely for the duration of the project and

then could be opened and the sediment could be returned to the excavated hole.

Beach and upper level sand should be stored in one geobag, while the deeper and

sub-tidal sands could be stored in a second geobag in order to sort the different

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sediment types. This would allow the sediments to be returned more precisely to their

original positions.

10.5 Harbour bed self-healing

Self-healing is a phrase used to describe the rate and final conditions to be

experienced at the construction sites after sediment is replaced or naturally infills

excavations. In many circumstances, particularly on aged sea beds, the action of

excavation has the effect of breaking the sediment bonds within the natural sands, by

allowing the loss of cohesive elements such as muds or biological agents. Once

disturbed, the sediment is more prone to erosion. The problem is particularly

pronounced in fine sediments where consolidation can take some time to occur, and

the bed in the meantime can have a quicksand feel.

The main area of concern for the southern route was the large inter-tidal banks on the

eastern side (Option 5C). However, this route is not recommended here. Instead, the

submarine pipe would traverse along the railway causeway, then across the channel

and through the two sand banks on either side near the existing railway causeway

(Option 5B). The grain sizes in this region are medium sands.

Along this route, it is ASR’s opinion that the “self-healing” will happen relatively quickly

(weeks to months, rather than years) because of the grain size and the effect of tidal

soakage through the sediments each tide. The sand will consolidate and there will be

no “quicksand” at the infilled excavations.

However, there will be a loss of some material during and immediately after the

excavation and so an allowance will need to be made to replace those volumes if the

original bed level is to be restored. The losses are likely to be up to 10-20% of the

amount excavated depending on the methods adopted during construction. This

sediment will have to be sourced.

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11 Greenhouse Sea-level Rise

In this section, we examine the impact of a 0.5 m “Greenhouse” sea level rise on the

circulation in the region of interest.

11.1 Methods

To generate boundary conditions, we simulated Tauranga Harbour using the same 75

m grid bathymetry, but with a 0.5 m sea level increase added to the initial tidal signal

applied at the northern open boundary. Boundary conditions for the nested 15 m cell

size sub-grid covering the main channel south of the Port of Tauranga to south of the

rail bridge (Figure 11.1) were extracted from the 75 m model simulation. Perpendicular

velocity data were extracted for application on the northern boundary of the sub-grid

and sea level data were extracted for the southern boundary.

This method is identical to that applied for previous simulations undertaken in this

study, with the addition of the 0.5 m sea level increase. The nested sub-grid includes

the recent “Walkway” bathymetric alterations representing a new walkway along the

western shoreline.

11.2 Results

General impacts of sea level rise could include increased coastal erosion and coastal

flooding, wetland/lowland inundation, impacts on coastal structures, increased salinity

of brackish and freshwater environments. Here we present the results of the model

simulations in the context of the present project and briefly describe the impacts.

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Results of the models simulations are shown in Figures 11.2-11.8. It is evident that:

• The tidal amplitudes increase slightly (Figures 11.2- 11.4)

• Currents are marginally stronger (about 5-10%), particularly over the tidal flats

(Figures 11.5–11.8)

• The existing residual circulation is strengthened, rather than changed in the

pattern (Figure 11.5).

Both of these outcomes could be anticipated for the following reasons. In the upper

reaches of a harbour, the tidal oscillation is delayed and reduced in amplitude by

frictional resistance along the flow path from the entrance. In deeper water (with the

sea level rise included), the effective friction is reduced and so the tidal amplitude

increases, i.e. water levels are higher at high tide and lower at low tide. The additional

volume flux associated with this goes in synchrony with a strengthening of the current

flows.

Model simulations do not show significant increase in current velocities near the

proposed western shoreline walkway (Figure .11.5 – 11.8). However, as the structures

have been designed for existing sea levels and conditions, the slightly faster currents

could exacerbate the predicted undermining of the toe of the walkway embankment.

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Figure 11.1. 15 m cell size nested model subgrid. This bathymetry includes a new walkway and pipeline configuration. Numbers indicate points were data was extracted for comparison below.

Figure 11.2. Sea level data at points 1-3 (top to bottom) for both the existing tidal situation and 100 yr sea level rise case.

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a

b

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Figure 11.3. Difference in sea levels at 3 points: (a) northern boundary - cell (75, 200), (b) under rail bridge – cell (53 55), and (c) southern boundary – cell (70, 1).

c

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Figure 11.4. Velocity comparison between current tidal state and 100 year sea level rise

estimate at the three points from above (a-c) by speed and direction.

a

b

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Figure 11.4. contd….

Figure 11.5. Difference in residual velocity vectors between normal sea levels and 100 year

sea level rise estimates.

c

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Figure 11.6. Residual velocity vector averages over a 15 day spring-neap cycle including a 100 year sea level rise of 0.5 m.

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Figure 11.7. 100 year sea level rise simulations showing maximum (a) flood and (b) ebb velocities throughout the 15 m cell size medium grid.

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Figure 11.8. 100 year sea level rise simulations showing maximum (a) flood and (b) ebb velocities near the rail bridge within the 15 m cell size medium grid.

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12 References

Ali, K.H.M and Karim, O., 2002. Simulation of flow around piers. Journal of Hydraulic Research, Vol. 40 (2), 161-174.

Black, K., 1983. Sediment transport and tidal inlet hydraulics. Doctor of Philosophy thesis. University of Waikato. New Zealand. Volume 1, Text, 331 pp. Volume 2, Figures and Tables

Black, K., 1984. Sediment Transport. Tauranga Harbour Bridge. Consulting Report to the Bay of Plenty Harbour Board, New Zealand. 112 pp.

Black, K.P., 1984. Sediment Transport. Tauranga Harbour Study. Consulting Report to the Bay of Plenty Harbour Board, New Zealand. Volume 1 (Text) 159 pp. and Volume 2 (Figures and Tables).

Black, K., 1987. A numerical sediment transport model for application to natural estuaries, harbours and rivers. In: Numerical Modelling: Applications to marine Systems, (Ed. J. Noye), Elsevier Science Publishers B.V. (North-Holland), 1987; 77-105

Black, K.P., Green, M., Healy, T., Bell, R., Oldman, J., and Hume, T., 1999. Lagrangian modelling techniques simulating wave and sediment dynamics determining sand bodies, in J. Harrf, W. Lemke, and K Stattegger (editors) Computerized Modeling of Sedimentary Systems, Springer, pp 3-21.

Black, K.P.; Bell, R.G.; Oldman, J.W.; Carter, G.S. and Hume, T.M., 2000. Features of 3- dimensional barotropic and baroclinic circulation in the Hauraki Gulf, New Zealand. New Zealand Journal of Marine and Freshwater Research , 34, 2000, pp 1 – 28

Black, K.P., 2001. The 3DD Computational Marine and Freshwater Laboratory. ASR Ltd. P.O. Box 67, Raglan. New Zealand.

Black, K.P. and B. Beamsley, 2003. Opureora (Matakana Ferry) Channel Dredging: Numerical Modelling. Report prepared for Boffa Miskell Limited on behalf of Environment Bay of Plenty, October 2003. pp60.

Breusers, H. N. C., 1965. Scour around drilling platforms. Bulletin, Hydraulic Research, 1964 and 1965, I.A.H.R., 19: 276-288.

Chase, K.J., and Holnbeck, S.R., 2004. Evaluation of pier-scour equations for coarse-bed streams. U.S. Department of the Interior, U.S. Geological Survey, Scientific Investigations Report 2004-5111, 18p.

Chiew, Y. M., 1984. Local scour at bridge piers. Report No. 355, University of Auckland, School of Engineering, Auckland, New Zealand.

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Dargahi, B., 1990. Controlling mechanism of local scouring. Journal of Hydraulic. Engineering, ASCE, Vol 116, No.10, pp. 1197-1214.

Demetrius, R.D., 2006. Bridge pier scour in tidal environments. Doctor of Philosophy Thesis, Department of Civil and Environmental Engineering, University of Maryland, USA, 896p

Engelund, F and Hansen, E., 1972. A Monograph on Sediment Transport in Alluvial Streams. Teknisk Forlag, Copenhagen.

Healy, T. (Editor) 2001. Challenges for the 21st Century in Coastal Science, Engineering and Environment. Journal of Coastal Research Special Issue 34, 687 p.

Healy, T., 1994: Channel dredging, dredge spoil migration, and downdrift impacts at a large tidal inlet, Tauranga Harbour, New Zealand. New Zealand Geographer 50(1),3-6.

Healy, T.,J. Mathew W. de Lange, and K. Black, 1997. Adjustments toward equilibrium of a large flood-tidal delta after a major dredging program, Tauranga Harbour, New Zealand. Proceedings of the 25th International Conference on Coastal Engineering, American Society of Civil Engineers, pp 3284-3294.

Healy, T.R. and R.M. Kirk, 1981(2nd edition 1992): "Coasts", Chapter 5 in J.Soons and M.J. Selby (eds.) Landforms of New Zealand. Longman-Paul. pp80-104.

Jelgersma, S, T. Healy, and E. Marone, 2002. Relative sea level changes and some effects on muddy coasts, in T. Healy, Y. Wang and J.A. Healy (Editors) Muddy Coasts of the World – Processes, Deposition and Function, . Elsevier Scientific, Amsterdam, pp 38-97

Johnson, P. A. and McCuen, R. H., 1991. A temporal spatial pier scour model. Transp. Res. Rec. 1319, Transportation Research Board, Washington D.C., 143-149.

Kruger, J., and Healy T., 2004. mapping the morphodynamic units of an ebb tidal delta: an application of the bottom-type classification for complex seafloors, Tauranga Harbour, New Zealand, Journal of Coastal Research (in press)

Mathew, J., T. Healy, M. Green, and K. Black, 1997. Investigation of sediment transport from an inner shelf dump ground, Tauranga, New Zealand. Second Indian National Conference on Harbour and Ocean Engineering (Inchoe-97), pp 1061-1072.

Mathew, J., T.R.Healy, and W. de Lange, 1995. Changes to a large flood tidal delta system following major channel dredging, Tauranga Harbour, New Zealand. Proceedings International Conference on Coastal and Port Engineering in Developing Countries (COPEDEC), pp 1465-1478.

Mathew, J.; Healy, T.R.; de Lange, W.P. and Immenga, D., 1997. Ebb-tidal delta response to shipping channel dredging, Tauranga, New Zealand. Pacific Coasts and Ports’97, 13th Australasian Coastal and Ocean Engineering Conference and

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6th Australasian Port and Harbour Conference, Christchurch, 7-11 September 1997, pp. 823-828.

Melville B.W., 1988. Scour at Bridges’, Chapter 15 of Civil Engineering Practice2 (Eds. Cheremisinoff, P.N. Cheremisinoff, N.P., and Cheng, S.L), Technomic Publishing Co., Pennsylvania, USA.

Melville, B. W. and Coleman, S. E., 2000. Bridge scour. Water Resources Publications, LLC, Littleton, Colorado.

Middleton, J.F. and Black, K.P., 1994. The low frequency circulation in and around Bass Strait: a numerical study. Continental Shelf Research 14(13/14): 1495-1521.

Raudkivi, A.J.,1986. Functional trends of scour at bridge piers. Journal of Hydraulic. Engineering, ASCE, Vol 112, No1, pp.1-12.

Richardson, E.V., and Davis, S.R. (1995). Evaluating scour at bridges. Report No. FHWA-IP-90-017, Hydraulic Engineering Circular No. 18, (3rd edition), Federal Highway Administration, Washington, DC.

Tian, F. 1997. Environmental aspects of storm runoff discharge from a timber port, Tauranga, New Zealand. Doctor of Philosophy thesis. University of Waikato. New Zealand, 279p.

Tian, F., A. Wilkins, and T. Healy, 1998. Accumulation of resin acids in sediments adjacent to a log handling area, Tauranga Harbour, New Zealand. Bulletin of Environmental Contamination and Toxicology, 60, 441-447

Tian, F., and Healy, T., 1998 Light attenuation by soluble and particulate substance in storm runoff from a timber port, Tauranga, New Zealand.. Journal of Marine Environmental Engineering, 4, 301-310.

Tian, F., and Healy, T., 1998 Optical properties of storm runoff from log handling areas at the Port of Tauranga, New Zealand. II. Colour. Journal of Marine Environmental Engineering 4, 311-329.

Tian, F., Healy, T.R.,and Wilkens, A., 1995. Aspects of runoff contamination from log handling areas at a timber export port, Tauranga, New Zealand. Journal of Marine Environmental Engineering, 1, 231-246.

Tian, F., Wilkins, A.L. and T.R. Healy. 1999. Extractives in storm runoff from a major timber port, Tauranga, New Zealand, Journal of Marine Environmental Engineering, 5, 85-105.

Young, I.R.; Black, K.P. and Heron, M.L., 1994. Circulation in the Ribbon Reef Region of the Great Barrier Reef. Continental Shelf Research, 14:2/3. pp. 117-142.

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13 Appendix 1 Velocity vectors at 1 hourly interval over 24 hours

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 208 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 209 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 210 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 211 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 212 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 213 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 214 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 215 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 216 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 217 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 218 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 219 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 220 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 221 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 222 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 223 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 224 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 226 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 227 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 228 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 229 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 230 hours

x (m)

y (m

)M

atapihi

Waipu Bay

ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 231 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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ASR Ltd Model 3DD

0.5 m/s

0.2 km

150 300 450 600 750 900 1050 1200 1350 1500

150

300

450

600

750

900

1050

1200Velocity vector at t = 232 hours

x (m)

y (m

)M

atapihi

Waipu Bay

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14 Appendix 2: Supplementary Studies

14.1 A2.1 Velocity Profiles

To determine the current forces on the dredging equipment at the sea bed, plots of the

velocity versus depth through the water column along the bridge crossing were provided

at 3 sites. The sites were specified by URS as shown in Figure A2.1.

Figure A2.1. Approximate locations for velocity profiles. Note that the locations of Points 2 and 3 have been switched for the profiles in Figure 12.

To generate the plots, the depth-averaged velocity generated by the numerical model

was used in the following formula which represents a logarithmic velocity to depth

relationship:

U = Umean log10(z / 0.002) / log10(0.37d / 0.002)

Where U is the velocity at elevation z above the seabed and Umean is the depth averaged

velocity produced by the numerical model. The roughness length is 0.002 m and d is the

water depth. We have provided plots for peak ebb and flood tide velocities for both

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spring and neap tides. These are shown in the Figure A2.2. On each figure we plot the

profile for a roughness length of both 0.002 m and 0.009 m to show the relatively small

influence of this parameter.

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a

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b

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c

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Figure A2.2.a–d. Velocity profiles for three locations for maximum ebb and flood flows at both spring and neap tide.

d

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14.2 A2.3 Discharge of Cement Casing Water

When pouring the concrete into the pier casings, the displaced water has contact with

the cement and requires disposal. The amount of displaced water is up to 100 m3 for

each of the 26 additional piers. The piers will be poured at around 1-2 per month, with a

maximum cement pour rate of 250 l/min. The water is to be stored in a ballast tank for

settlement of the solids and then discharged to the harbour. The pH of the water is up to

12 and so the water needs to be discharged slowly through a diffuser to achieve

dilutions of at least 200:1.

The water could be discharged slowly through a diffuser attached vertically to an existing

bridge pile. In the deepest water and with some allowance for tides, it was estimated that

the diffuser length could be up to 5 m.

If the plume mixes into the water column next to the pile, it would therefore discharge

into a region some 5 m high and some 0.5 m wide. The discharge rate over 1 day would

be

Q= 100 m3/(1 day*24 hr/day * 3600 sec/hr) = 0.00116 m3/s

Table A2 shows the dilutions for various flow rates past the pile when discharged into a

region some 5 m x 0.5 m in area. The dilution is substantial. Indeed, as long as only 1

pier is drilled every week, and discharge is 0.0016 m3/s, the total 100 m3 can be

discharged on the ebb tide only over a 6 day period for 4 hours of the ebb tide twice a

day.

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Table A2.2: Dilution calculations for concrete laden water.

Tauranga City Council Southern Pipeline

Dilution of the discharge of pier casing water displaced during cement pours

Volume of cement water displaced per pier 100 m3

Rate of cement pour 5.5 L/s

Dilution required 200:1

Discharge period (assumed) 1 days

Diffuser length 5 m

Width of plume discharge

0.5 m, assuming this is the width of the pier the flow is discharged across (ie at the 'bow' of the pier section)

Discharge Q= volume/# days 0.00116 m3/s

Dilutions for various harbour flow rates

Harbour flow velocity (m/s)

Volume flow rate of sea water passing (m3/s)

Volume flow rate of effluent (m3/s)

Dilution Ratio :1

Satisfactory Dilution?

Number of days to discharge if released over 4 hour peak ebb tides only

0.1 0.25 0.00116 216 OK

0.5 1.25 0.00116 1080 OK

0.9 2.25 0.00116 1944 OK

6

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15 Appendix 3: Analysis of shoreline change

This appendix provides the technical information for the analysis of the shoreline

changes in the 3 aerial photographs shown in Figure 2.28.

Software used:

U14_5_3.sid U14_4_2.sid U14_4_3.sid U14_5_2.sid Images

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Image 2002 harbour crossing added

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Locations that appear on the Base image and 2002 harbour crossing added are used for ground control points. Green crosses are on 2002 harbour crossing and Yellow crosses on the Base image.

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Locations are chosen that are; as obvious as possible in both images, cover as much of the area of the image as possible, are very unlikely to have moved, are close to the ground and not prone to perspective distortions. Locations on the inter tidal zone are difficult to consolidate, only one has been chosen (top right). The area close to the bridge has a greater number of points to aid rectification in this important area.

Image pixel location X Y map coordinates Easting Nothing

1086.377968 -44.411616 2789873.920190 6386061.314313

1057.804467 -189.097438 2789844.827170 6385913.511382

949.168655 -666.627970 2789731.572200 6385418.930044

1408.420770 -587.523366 2790206.025225 6385498.647158

2713.879967 -886.050974 2791550.324728 6385188.036524

2730.195607 -976.421530 2791567.600093 6385096.045203

2963.789067 -1830.245900 2791806.259751 6384214.623752

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2904.091658 -2019.262576 2791746.912824 6384020.435964

2758.199896 -1861.468373 2791595.256353 6384182.553348

18.056669 -1934.890920 2788769.594897 6384110.985929

24.036776 -1728.332354 2788774.776928 6384324.246452

137.983064 -1308.141234 2788893.580868 6384760.061742

574.832427 -68.703435 2789344.787972 6386038.145975

342.797975 -550.242281 2789105.196576 6385539.429366

568.071891 -1486.356306 2789333.294324 6384576.488491

2021.246404 -1271.685999 2790835.861222 6384794.619102

2308.004385 -726.679204 2791132.907224 6385355.358247

756.389156 -925.609898 2789528.994414 6385152.744266

2524.304601 -668.311021 2791357.283404 6385414.064638

584.350317 -1849.915912 2789352.310964 6384197.147817

175.877644 -789.685247 2788933.474677 6385292.915084

2231.694606 -1647.360259 2791053.629450 6384402.835851

802.317443 -539.908609 2789578.746383 6385549.994982

887.368658 -445.270451 2789666.766083 6385651.813995

903.239794 -459.635517 2789684.324138 6385633.392429

1015.303280 -583.746036 2789798.925897 6385503.300776

961.263787 -247.126890 2789743.229442 6385854.119681

775.689554 -124.865266 2789554.446835 6385984.727792

683.093069 -73.120854 2789457.187339 6386032.891012

1001.243912 -423.714948 2789785.886047 6385670.697138

2887.203912 -150.757301 2791731.453364 6385948.984094

459.486300 -2084.991035 2789223.234234 6383956.047727

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RMS 1.86922

1st order polynomial transformation. 1978 aerial georesited to the base Image

Locations that appear on the Base image and 1978 image added are used for ground control points. Green crosses are on 1978 image and Yellow crosses on the Base image. Determining locations that match the base image prove difficult. Roof tops are the common GCP feature. The bridge and the wharves have also been determined to be visible and continual. Locations on the inter tidal zone are difficult to consolidate, none were used

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Image pixel location X Y map coordinates Easting Nothing

379.181875 -638.283305 2789874.068030 6386062.484576

752.232769 -1292.808415 2790203.676160 6385494.844995

126.420585 -1456.082412 2789664.576964 6385349.247054

264.161638 -1282.729031 2789776.276365 6385499.541777

136.599561 -1360.544982 2789674.566341 6385430.978323

2211.269994 -2381.159970 2791483.269208 6384551.355918

1961.822612 -1109.262913 2791260.863720 6385662.429641

383.609553 -328.284824 2789876.233913 6386331.708533

150.576736 -147.673123 2789679.402949 6386479.233483

1750.143052 -1849.224264 2791078.130251 6385015.781104

306.451240 -1049.771900 2789814.246421 6385700.953917

1727.235314 -2576.482011 2791062.847502 6384381.180516

1516.960638 -1615.413343 2790873.061917 6385216.392263

RMS for 1978 is 3.70130

1st order polynomial transformation

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1947 image

Locations that appear on the Base image and 1947 image added are used for ground control points. Green crosses are on 1947 image and Yellow crosses on the Base image. Determining locations that match the base image prove difficult. Roof tops are the common GCP feature. The bride also was determined to be visible and continual. Locations on the inter tidal zone are difficult to consolidate, none were used.

Image pixel location X Y map coordinates Easting Nothing

1501.601454 -353.219364 2790206.696064 6385498.814741

380.624270 -1175.720017 2789296.997954 6384880.933647

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589.147244 -978.159386 2789467.924988 6385033.269107

1014.303925 -149.754980 2789822.080443 6385671.144299

1253.617820 -295.122149 2790013.709151 6385550.996141

1102.445227 -238.943066 2789888.744927 6385599.093313

186.535868 -1540.987749 2789137.858598 6384599.743814

977.733643 -340.812283 2789788.117929 6385522.348553

833.835961 -600.341226 2789668.231142 6385320.700110

824.462943 -700.868847 2789660.508301 6385242.399077

343.529304 -766.813246 2789277.211023 6385203.381563

RMS for 1947 1.66285

15.1 A3.1 Interpreting the root mean square error

When the general formula is derived and applied to the control point, a measure of the

error—the residual error—is returned. The error is the difference between where the

from point ended up as opposed to the actual location that was specified—the to point

position. The total error is computed by taking the root mean square (RMS) sum of all

the residuals to compute the RMS error. This value describes how consistent the

transformation is between the different control points (links). When the error is

particularly large, you may want to remove and add control points to adjust the error.

Although the RMS error is a good assessment of the accuracy of the transformation,

don’t confuse a low RMS error with an accurate registration. For example, the

transformation may still contain significant errors due to a poorly entered control point.

The more control points of equal quality used, the more accurately the polynomial can

convert the input data to output coordinates. Typically, the adjust and spline

transformations give an RMS of near zero or zero; however, this does not mean that the

image will be perfectly georeferenced.

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1947 digitizing of shoreline. Gaps in line are where shadows prevent any certainty of the location of the shore.

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“Light” area on shore determined to be the shoreline. Tidal levels not available , water clarity issues.

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1978 digitizing of shoreline. Gaps in line are where shadows prevent any certainty of the location of the shore.

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“Light” area on shore determined to be the shore line.

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Two lines are used to show the apparent shoreline (RH side) and the known seawall (LH side).

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2002 digitizing of shoreline

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